The TRB analysed in this paper appertains to a building typology highly recurrent in Eastern Sicily (Italy) and devoted to wine storage and production. It is a one-storey TRB (Figure 2) located at an altitude of 350 m above sea level in the Etnean area which is one of the most earthquake-prone Italian areas. To date, it is used as warehouse for agricultural products. The walls are made of well-structured rough-shaped basaltic stone masonry which is highly frequent in the considered area (28). The double-pitched roof has a wooden bearing structure and is covered by traditional curved tiles made of baked-clay.

Some alterations of the original character of the facades, probably due to a variation of its original intended use, resulted in the misalignment of the external wall openings. The lintels of the openings are built by using materials having different mechanical characteristics. This is the case of many TRBs located in the Etnean area, where previous studies revealed that openings have an internal and an external flat-arch lintel: the external one is made of two blocks of lava-stone, the internal one is made of wooden beams (28).

In this paper a further geometrical configuration of the wall openings was considered by assuming the building to be reused for rural tourism activities (e.g., tourist accommodation and agriculturally-based activities). This involved the lowering of some window openings down in order to assure a view from the inner rooms to the outside, and widening some door openings to allow vehicle passage.

In this study, the structural scheme used to simulate the behaviour of the building under study considered each wall independently, as set out in Section C8.7.1.1 of DEN09 and adopted in other research (29). In detail, the building was subdivided into four perimeter walls which were separately studied, by modelling each shear masonry wall with openings as a shear-type structural system. In such a system the wall is composed of pier panels and masonry spandrels.

By applying the seismic safety assessment model described in a previous work (17) and summarized in Figure 3, the in-plane structural behaviour of the East longitudinal wall, analyzed in the current functional destination (Figure 4a) and in the hypothesized adaptive reuse (Figure 4b), is discussed in the following of this paper. This longitudinal wall was selected among the others as it had the lowest wall-to-opening ratio, and showed the absence of connected crossing walls which would have improved its overall resistance.

The mechanical resistance parameters of this masonry, classified as ‘well-structured rough-shaped stone masonry’, are reported in Table 1 according to the considered standards. The values of the masonry compression resistance (f_m) in NTC08 are comparable with those adopted in studies on stone masonry (30) or resulting from tests using flat jacks (31). Studies on lava-stone masonry (32) (33) recommend a masonry specific weight value (γ _m) in the range 18 ÷ 25 10^–6 Nmm^–3. With regard to these parameters, it should be underlined that many factors affect masonry in-plane behaviour, such as the geometry of the wall (i.e., height, width, thickness), the mechanical and geometrical properties of the masonry constituents (stones and mortar), the building techniques adopted, the geometrical characteristics of the openings.

As uncertainty is typical of structural evaluation of existing building, EC8-3 and NTC08 introduced different sets of materials, structural safety factors and associated analysis procedures, defining three knowledge levels (KLs) related to the mechanical properties of the masonry constituents, and the geometrical properties and building details (i.e. connection type and reinforcing components) of the structure to be evaluated. In the absence of experimental values of resistance for masonry walls, the knowledge level could be KL1 or KL2 type (namely ‘limited’ and ‘normal’ knowledge level in EC8-3). In Table 1the confidence factors (CF) related to the KLs were reported.

Tri-linear characteristic (i.e., force-displacement) curves of the pier panels were constructed by using a non-linear algorithm applied in previous studies (17) (34). In the adopted model (Figure 3), the principal failure mechanisms of masonry pier panels due to shear and flexure were considered: diagonal-cracking shear failure, bed-joint-sliding shear failure, and flexural failure. For the analysed masonry wall, the shear capacities corresponding to diagonal-cracking shear collapse (V_u,diag) and bed-joint-sliding shear collapse (V_u,oriz) were calculated in compliance with NTC08.

Since the use of different relations to compute the ultimate capacities V_u of the masonry pier panels may influence the overall performance of the wall, in this paper the flexural capacity V_u,flex was calculated in compliance with NTC08 (V_u,flexNTC) as well as according to some Authors (34) (V_u,flexAU). For the computation of V_u, in the absence of specific experimental data, some masonry strength characteristic values (e.g., flexural resistance of the spandrel) were conservatively set according to Fusier and Vignoli (34). It is widely recognized, in fact, that in this field there are few experimental studies due to the difficulty in obtaining these values (35).

With regard to the computational height of the pier panels (h), an univocal relation was not found in literature (35) (36) and in specific standards. In Italy a specific technical norm (37) suggests that h should be “equal to the interstorey height though, strictly speaking, it would be more correct to consider the height of the openings”. Therefore, the model was applied in order to evaluate wall behaviour in relation to three different scenarios of pier-panel height: the first is the height of the openings (hp₁); the second is a function of the ‘effective height’ as reported by Dolce (38) (hp₂); the third (hp₃) is the distance between the wall base and the axes of masonry spandrels.

The graphical symbols for the collapse mechanisms of pier panels and masonry spandrels were reported in Figure 5. Here, condition δ₁ ≤δ ≤δ₂ refers to the displacement related to the elasto-plastic phase of the characteristic curve of the pier panel, whereas condition δ₂ ≤δ ≤δ₃ indicates the displacement related to the plastic phase (17) (34).

The conventional seismic loads (F_s) for the analysed masonry wall were computed by applying the two static methods described in NTC08 and NTC05. In general, the limit state of reference for masonry structures is the Ultimate Limit State (ULS). In NTC05 the ULS associated to a reference probability of exceedance (RPE) of 10 % in 50 years, corresponds to the significant damage state (SD) of EC8 and to the life safety limit state (LSLS) of NTC08. The collapse limit state (CLS) defined in NTC08 refers to a RPE of 5 % in 50 years. Moreover, the relation for computing the seismic action according to NTC05 is straight forwardly derived from EC8, whereas its definition in NTC08 for that action is site specific depending on the location of the buildings. Therefore, in the analysed case study the coefficients required for the F_s computation were related to the second category seismic areas (medium seismicity) as per NTC05, whereas for CLS and LSLS they were derived from tab.1-Annex B of NTC08 that provides values related to TRB’s geographical coordinates.

The force-based incremental iterative procedure shown in Figure 3 allowed the computation of the capacity curve of the masonry shear wall as well as the safety verification of masonry spandrels.

Exhaustive experimental trials are necessary to find out the real flexural strength capacity of masonry spandrels of existing buildings, taking into account their different typologies. The literature lacks experimental studies which can thoroughly evaluate resistance, strain behaviour and dissipative capacity of masonry spandrels subject to seismic forces (33).

EC8-1 states that: “In the structural model masonry spandrels may be taken into account as coupling beams between two wall elements, if they are regularly bonded to the adjoining walls and connected both to the floor tie beam and to the lintel below”. The NTC05 considers, instead, that a certain flexural resistance could develop in masonry spandrels if there is at least one horizontal tensile-resistant element (e.g., chains, lintels, tie beams, steel ties, and other strengthening elements like fibre reinforced polymer bands). When an axial force comes about thanks to these other structural elements, masonry spandrels can be simulated as:

Figures 6a, Figure 6b, and 6c show the results of the seismic assessment simulations to evaluate the wall behaviour in relation to the different scenarios of pier-panel height and masonry spandrel modelling. In detail, the figures report the crack development of pier panels and masonry spandrels at the increasing of the horizontal in-plane forces (F_h), which are expressed as a function of F_s, up to wall collapse. Moreover, spandrel cracking is depicted in thin black lines for SM1 and in thick black lines for SM2. The graphical symbols related to the collapse mechanisms reported in Figure 6 are clarified in Figure 5.

In each figure, the bottom scheme shows the first pier-panel cracking, meaning that the elastic phase of the panel was over and the elasto-plastic phase had begun, and/or the first failure in masonry spandrels. The top scheme shows wall collapse at ultimate in-plane strength capacity (F_u) and the ultimate displacement of the i-th pier panel (δ_u(i)). The other schemes show intermediate damage levels, e.g., the elasto-plastic limit of all the pier panels, when the first cracking takes place in all the pier panels; plasticity, when one of the pier panels reaches the plastic phase of its characteristic curve; seismic performance state, when F_h about equals the seismic force F_s; the shear or flexural cracking of all the masonry spandrels when the shear or flexural verification of all masonry spandrels fails.

The results of the simulations as per NTC08 showed some differences compared to those obtained using NTC05. Though the crack development of pier panels and masonry spandrels was similar yet F_u varied since some pier panels exhibited different collapse mechanisms and thus their characteristic curves differed. In the scenario hp₃, for instance, the collapse mechanism of the first pier panel changed from flexural failure as per NTC05 to bed-joint-sliding shear failure according to NTC08. Moreover, the ULS of NTC05 always led to lower wall seismic safety coefficients ν (defined as the ratio between F_u and F_s) than the LSLS of NTC08 while always producing higher ones than the CLS of NTC08.

The results showed that the simulations were affected by the application of both the different values of the mechanical characteristic recommended by the two standards and the different computation of F_s as per the two standards. In fact, according to NTC08 the values of f_m and the G modulus increased nearly twofold if compared to those recommended by NTC05 (Table 1). As regards the computation of F_s, the NTC08 gave more geographically differentiated values for seismic load as its computation parameters were related to geographic coordinates rather than to uniform values in each of the seismic zones defined by NTC05.

The KL variation from KL2 to KL1 highlighted once more a different mechanical behaviour for some of the pier panels. According to NTC05, the reduction of ν due to the considered KL variation was lower than that obtained as per NTC08. This was caused by the different values of the elastic modulus provided by the standards (Table 1). Moreover, the increase in CF values due to the variation of KL produced a reduction in strength capacity causing a change in pier-panel collapse mechanisms. Therefore, the building characterization of TRBs as well as the analysis of their current conservation state are crucial to define the appropriate CF value in the norms, especially if the building had been subject to different rehabilitation works over the years (25).

By considering the wall seismic safety coefficients ν, Figure 6a, Figure 6b, and Figure 6c show that the scenario hp₁ was less conservative than the others since it provided higher strengths and safety coefficients, while the scenario hp₂ led to intermediate results of hp₁ and hp₃.

Furthermore, the different scenarios of h resulted in different wall damage conditions. In fact, the first cracking of masonry occurred in different pier panels: the sixth one in hp₁, the third one in hp₂ and the seventh one in hp₃. When the strength capacity of the wall exceeded F_s, the scenario hp₁ always determined pier panels to reach the plastic phase for F_h/F_s values higher than those related to hp₂ and hp₃ (for instance, in the LSLS state the F_h/F_s values were 1.80, 1.65, and 1.49 in the three scenarios of h, respectively, as shown in Figure 6a, Figure 6b, and Figure 6c).

These results are also interrelated to pier-panel collapse mechanisms which in turn depended on h. In fact, the collapse mechanisms of almost all the pier panels changed from diagonal cracking failure in the scenario hp₁ to bed-joint-sliding shear failure and flexural failure in hp₂ and hp₃.

By analysing the pier-panel characteristic curves corresponding to the three different scenarios of h, the most frequent pier-panel failure mechanism was diagonal cracking shear failure though there was also some bed-joint-sliding shear failure. In general, for low constructions with wide pier panels, the most frequent condition determining collapse is diagonal cracking shear failure. Yet, when normal stress decreases, the failure is not due to exceeding the tensile resistance governing the collapse mechanism of shear with diagonal cracking, but rather by exceeding the bed-joint-sliding shear resistance (Figure 7a). These changes in collapse mechanisms affected wall collapse as they caused different pier panel to reach failure (top schemes in Figure 6a, Figure 6b, and Figure 6c). Yet, collapse mechanisms depended also on pier-panel slenderness. In fact, when analyzing the variation of pier-panel failure mechanisms in relation to both their slenderness and normal stress (σ₀), it resulted that:

Therefore, in the simulations it was found that, when σ₀ kept under the threshold value, failure varied from diagonal cracking shear failure to bed-joint-sliding shear failure at increasing of the slenderness. For instance, this occurred for the second pier panel. When pier panels were simulated as slender (i.e., slenderness ≥ 1.5), for instance the third one in hp₃, failure was due to flexure whereas diagonal cracking shear failure and bed-joint-sliding shear failure were the failure modes in hp₁ and hp₂, respectively (Figures 6a, Figure 6b, and Figure 6c). These different predictions of failure mechanisms would modify the pier-panel characteristic curves as well as the capacity curve of the whole structure.

Moreover, when different relations were used to account for the flexural capacity of masonry pier panels (34) (41), results showed that in some cases the condition V_u,flexAU<V_u,oriz occurred for low axial-force values typical of masonry buildings and NTC08 overestimates the ultimate force (V_u,flexNTC) (Figure 7a), resulting in wrong predictions of failure mechanisms for some values of the applied axial force (41). Also this overestimation could affect pier-panel characteristic curves as well as the capacity curve of the whole structure.

In SM1, diagonal-cracking shear failure for well-sized spandrels, was more limiting than other failure mechanisms, which in turn would affect the spandrels having a higher length/height ratio (r). In fact, according to SM1, all the cases of spandrel failure due to shear occurred before flexural failure (Figures 6a, Figure 6b, and Figure 6c). Generally, the spandrels having a more uniform r (fixed shear force transmitted by the masonry piers, fixed h), show failure mainly due to shear while spandrels having a higher and more differentiated r frequently exhibit flexural failure (data not reported). Under SM2, flexural failure did not occur in the examined cases.

The results related to the reuse simulation (Figure 6d), computed for hp₃ which showed the most diffuse spandrel cracking, demonstrated that the considered geometrical modifications of wall openings position and width improved both pier panel and masonry spandrel resistances. In fact, the first cracking and the beginning of the plastic phase of the pier panels occurred earlier in the simulation of the original building condition than in the reuse one, and also masonry spandrels cracking occurred later in the reuse simulation (Figures 6c and Figure 6d). Therefore, enhancing the horizontal regularity of the wall openings improved the overall wall resistance. This result confirmed what observed in previous research (18) (42).

Masonry spandrels, however, often exhibit shear failure and thus need to be strengthened in order to make it possible to use the shear-type modelling. By applying the model in the case of a masonry wall strengthened by mortar injections, a higher increase of ν and a higher resistance of spandrels were achieved in comparison to the case of unstrengthened masonry. In the LSLS limit state, for instance, ν increased from 1.77 (Figure 6d) to 2.07 and spandrel cracking was prevented for F=F_s in the SM2 case.

On the basis of these results it appears to be very important to perform experimental tests and refine models to achieve simulations of masonry building responses to earthquakes closer to their real behaviour, also taking into account the characteristics peculiar to specific typologies of buildings such as those considered in this study (25) (43) (44) (45).